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HomeMy WebLinkAbout13. DRAFT Geotech RptDRAFT GEOTECHNICAL REPORT LAKE TO SOUND TRAIL, BLACK RIVER BRIDGE RENTON, WASHINGTON HWA Project No. 2010-100 T200 February 24, 2015 Prepared for: Parametrix, Inc. DRAFT February 24, 2015 HWA Project No. 2010-100 T200 Parametrix, Inc. 719 2nd Avenue, Suite 200 Seattle Washington 98104 Attention: Ms. Jenny Bailey Subject: DRAFT GEOTECHNICAL REPORT LAKE TO SOUND TRAIL, BLACK RIVER BRIDGE RENTON, WASHINGTON Dear Jenny: Enclosed is our draft geotechnical report for the proposed Lake to Sound Trail, Black River Bridge in Renton, Washington. To stabilize the river banks during a design earthquake event per AASHTO LRFD bridge design specifications, stone column treatment is recommended. Deep pile foundations will also be necessary for bridge support. We appreciate the opportunity of providing geotechnical services on this project. We look forward to receiving your review comments on this draft report. Should you have any questions please do not hesitate to call. Sincerely, HWA GEOSCIENCES INC. Sa H. Hong, P.E. Principal Geotechnical Engineer DRAFT TABLE OF CONTENTS Page 1.INTRODUCTION .......................................................................................................1 1.1 PROJECT DESCRIPTION ...................................................................................1 1.2 SCOPE OF SERVICES AND AUTHORIZATION .....................................................1 2.FIELD AND LABORATORY INVESTIGATIONS ..................................................1 2.1 FIELD EXPLORATIONS .....................................................................................1 2.2 LABORATORY TESTING ...................................................................................2 3.SITE CONDITIONS ...................................................................................................2 3.1 SURFACE CONDITIONS ....................................................................................2 3.2 GENERAL GEOLOGIC CONDITIONS ..................................................................2 3.3 SUBSURFACE CONDITIONS ..............................................................................3 3.3.1 Soil Stratigraphy ............................................................................3 3.3.2 Ground Water .................................................................................4 4.CONCLUSIONS AND RECOMMENDATIONS ......................................................5 4.1 SEISMIC DESIGN .............................................................................................5 4.1.1 General ...........................................................................................5 4.1.2 Regional Seismicity .......................................................................5 4.1.3 Seismic Considerations ..................................................................6 4.1.4 Soil Liquefaction ............................................................................6 4.1.5 Ground Fault Hazard .......................................................................7 4.2 SLOPE STABILITY EVALUATIONS .....................................................................7 4.2.1 Static Slope Stability Analyses .......................................................7 4.2.2 Pseudo-Static Slope Stability Analyses ..........................................8 4.2.3 Post Liquefaction Slope Stability Analyses ....................................8 4.2.4 Lateral Spreading and Sliding .........................................................8 4.2.5 GLOBAL STABILITY AFTER GROUND IMPROVEMENT ....................................9 Static Slope Stability Analyses ................................................................9 Pseudo-Static Slope Stability Analyses ...................................................9 4.3 Ground Improvement Techniques (GIT) ...........................................10 4.5 BRIDGE FOUNDATIONS ..................................................................................11 4.5.1 Axial Loading .................................................................................12 4.6 Axial Resistance for Pile Design. ......................................................12 4.7 Laterally Loaded Driven Piles ............................................................14 4.8 DRIVEN STEEL PIPE PILE INSTALLATION AND CONSTRUCTION CONSIDERATIONS16 4.9 BRIDGE ABUTMENTS AND WING WALLS .........................................................16 4.9.1 Lateral Earth Pressures - Static Condition .....................................16 4.9.2 Lateral Earth Pressures during Seismic Loading ............................17 4.9.3 Abutment Wall Backfill ..................................................................18 4.10 SPREAD FOOTING BEARING CAPACITY ON EXISTING DENSE FILL .................18 DRAFT February 24, 2015 HWA Project No. 2010-100 T200 4.11 SLIDING RESISTANCE ON EXISTING FILL FOR CAST-IN-PLACE CONCRETE FOOTINGS ......................................................................................................18 4.12 STRUCTURAL FILL MATERIALS AND COMPACTION ........................................18 4.13 WET WEATHER EARTHWORK ........................................................................19 4.14 EMBANKMENT SLOPES ................................................................................20 4.15 SITE DRAINAGE AND EROSION .....................................................................20 4.15.1 Surface Water Control .................................................................20 4.15.2 Erosion Control ............................................................................21 5.CONDITIONS AND LIMITATIONS .........................................................................21 6.REFERENCES ............................................................................................................23 LIST OF TABLES Table 1. ........................................................................................................................6 Table 2. ........................................................................................................................9 Table 3. ........................................................................................................................10 Table 4. ........................................................................................................................12 Table 5. ........................................................................................................................13 Table 6. ........................................................................................................................13 Table 7. ........................................................................................................................14 LIST OF FIGURES (FOLLOWING TEXT) Figure 1. Vicinity Map Figure 2. Overall Site and Exploration Plan Figure 3. Bridge Area Site and Geotechnical Profile A-A APPENDICES Appendix A: Field Exploration Figure A-1. Legend of Terms and Symbols Used on Exploration Logs Figures A-2 – A-3. Logs of Boreholes BH-1 and BH-2 Appendix B: Laboratory Testing Figures B-1 – B-4. Particle Size Distribution Test Results Appendix C: Slope Stability Analyses Results 2010-100 T200 DR 2 HWA GEOSCIENCES INC. DRAFT DRAFT GEOTECHNICAL REPORT LAKE TO SOUND TRAIL, BLACK RIVER BRIDGE RENTON, WASHINGTON 1.INTRODUCTION 1.1 PROJECT DESCRIPTION HWA GeoSciences Inc. (HWA) completed a geotechnical study for the proposed Lake to Sound Trail Segment A, Black River Bridge in Renton, Washington. The location of the site and the general project layout are shown on the Vicinity Map (Figure 1) and the Site and Exploration Plan (Figure 2), respectively. The purpose of this geotechnical study was to explore and evaluate surface and subsurface conditions at the site and, based on the conditions encountered, provide recommendations pertaining to geotechnical aspects of the project. According to current design plans, the new trail pedestrian bridge will consist of a, single-span steel or concrete girder structure with a span of approximately 114 feet over the Black River. The bridge foundation will be supported on either drilled shaft foundations or driven steel pipe piles. The new bridge is being designed in accordance with AASHTO Load and Resistance Factor Design (LRFD) methodology. We understand wetland impacts will be mitigated to protect the wetland located north of the trail alignment, as well as the Black River channel. Geotechnical explorations were performed at the proposed ends of the bridge span to evaluate site soil and ground water conditions. 1.2 SCOPE OF SERVICES AND AUTHORIZATION Geotechnical engineering services was authorized in a subconsultant agreement dated November 7, 2014. Our scope of work included collecting and reviewing readily available geotechnical and geologic information for the area in the vicinity of the project site; coordinating the field activities with the project team; advancing two exploratory borings; performing laboratory testing and engineering analyses to develop geotechnical recommendations for the proposed improvements; and preparing a draft geotechnical report. 2.FIELD AND LABORATORY INVESTIGATIONS 2.1 FIELD EXPLORATIONS Two geotechnical explorations were conducted on November 10, 2014 and January 6, 2015. Borehole BH-1 was drilled at the north side of the river and BH-2 was drilled on the south side, with hollow-stem auger drilling methods. The explorations were supervised and logged by an DRAFT February 24, 2015 HWA Project No. 2010-100 T200 HWA geologist, who observed the exploratory work on a full time basis. A detailed discussion of the field exploration methodologies and the equipment used is presented in Appendix A, along with the borehole logs and a legend of terms and symbols used on the logs. The exploration locations are shown on Figures 2 and 3. 2.2 LABORATORY TESTING Laboratory tests were conducted on selected samples obtained from the borings to characterize relevant engineering and index properties of the site soils. Laboratory tests included determination of in-situ moisture content, and grain size characteristics. The tests were conducted in general accordance with appropriate American Society of Testing and Materials (ASTM) standards. The test results and a discussion of laboratory test methodology are presented in Appendix B, and/or displayed on the exploration logs in Appendix A, as appropriate. 3.SITE CONDITIONS 3.1 SURFACE CONDITIONS The proposed bridge alignment is located approximately 80 feet (south end) to 230 feet (north end) east of Monster Road Bridge in the City of Renton. The river banks are inclined at approximately 2H:1V. We understand the trail bridge approach will be slightly above the original ground surface on an embankment. Both banks are armored with rip-rap rock with a diameter ranging from 12 to 24 inches. 3.2 GENERAL GEOLOGIC CONDITIONS The geology of the Puget Sound region includes a thick sequence of glacial and non-glacial soils overlying bedrock. Glacial deposits were formed by ice originating in the mountains of British Columbia (Cordilleran Ice Sheet) and from alpine glaciers which descended from the Olympic and Cascade Mountains. These ice sheets invaded the Puget Lowland at least four times during the early to late Pleistocene Epoch (approximately 150,000 to 10,000 years before present). The southern extent of these glacial advances was near Olympia, Washington. During periods between these glacial advances and after the last glaciation, portions of the Puget Lowland filled with alluvial sediments deposited by rivers draining the western slopes of the Cascades and the eastern slopes of the Olympics. The most recent glacial advance, the Fraser Glaciation, included the Vashon Stade, during which the Puget Lobe of the Cordilleran Ice Sheet advanced and retreated through the Puget Sound Basin. Existing topography, surficial geology and hydrogeology in the project area were heavily influenced by the advance and retreat of the ice sheet. 2010-100 T200 DR 2 HWA GEOSCIENCES INC. DRAFT February 24, 2015 HWA Project No. 2010-100 T200 Surficial geological information for the site area was obtained partly from the published maps, “Geologic Map of the Renton Quadrangle, King County, Washington” (Mullineaux, 1965) and “Geologic Map of the Des Moines Quadrangle, King County, Washington.” (Booth and Waldron, 2004). The maps indicate that the uplands to the southwest and immediate north consist of Tertiary igneous bedrock predominantly mantled by Pleistocene Vashon till, while the valley floor is covered by alluvial deposits. The bedrock consists of highly jointed and faulted andesite. The till was deposited as a discontinuous mantle of ground moraine beneath glacial ice on the eroded surface of older deposits. Soils defined as Vashon till consist of an unsorted, non-stratified mass of silt, gravel, and sand in varied proportions. The till is of high density/strength due to glacial over-consolidation, and typically has low permeability. The 1965 map, which includes the subject site, indicates the valley floor is covered by alluvium deposited by the White River and Green River, prior to historical diversion of the White River south into the Puyallup in 1906. According to the map this alluvium consists of silt and fine sand at the surface, becoming medium to coarse sand with depth. Black volcanic sand is typical of White River deposits in the valley. The Black River formerly was the outlet for Lake Washington, prior to completion of the Lake Washington Ship Canal in 1917. Very little to no sediment would be expected to exit a body of water, and therefore Black River deposits would consist merely of reworked sediment of the Cedar River and White River. 3.3 SUBSURFACE CONDITIONS 3.3.1 Soil Stratigraphy Our interpretations of subsurface conditions were based on the results of field exploration, our review of available geologic and geotechnical data, and our general experience in similar geologic settings. It should be noted that in-situ tests performed during drilling, e.g. Standard Penetration Tests represented by N values, identified liquefiable fine sandy silt layers within both borings. For reference, the blow count values recorded during tests are included on the boring logs and are plotted on the penetration resistance chart on each log. Soil density descriptions on the boring logs are based on our observations of soil granularity vs. cohesiveness in addition to the recorded penetration values. In general, the area of the proposed bridge site is underlain by a sequence of layers of recent silt and sand alluvium deposited by the historical White River and Black River, underlain by either bedrock or glacial till. Suitable bearing material for bridge foundations was encountered at approximately 45 feet on the north bank (glacial till, over bedrock in BH-1) and at 67 feet at the south bank (glacial till in BH-2). The soil units encountered in the borings are described separately and in more detail below. The conditions are also summarized in Figure 3. Appendix 2010-100 T200 DR 3 HWA GEOSCIENCES INC. DRAFT February 24, 2015 HWA Project No. 2010-100 T200 A contains detailed summary logs of subsurface conditions encountered at the individual exploration locations. •Fill - Both borings encountered fill at the ground surface to depths of 7.5 feet in BH-1 and approximately 25 feet in BH-2. The fill consisted of medium dense to dense, gravelly silty sand in the upper 4 to 7 feet, then medium dense to loose sandy silt to silty sand with variable gravel content. In BH-2 this latter material had the appearance of alluvium with fine bedding below 17.5 feet, however a chunk of rubber in the sampler obtained from 20 feet indicated the material was fill to approximately 25 feet. Based on this depth of fill, we speculate that it originated as dredge tailings fill from channel modifications to the Black River. The protective surficial layer of fill on both banks of the river consisted of loosely placed riprap rocks. •Loose Alluvium - Recent alluvial deposits were encountered beneath the existing fill in both borings. The upper portion of alluvium in BH-1 consisted of fine sandy silt and silty sand. It was very loose with N values ranging from 0 to 5 and extended from approximately 7.5 to 30 feet deep. In BH-2, loose alluvium consisting of slightly silty sand and sandy gravel was encountered from 25 to 40 feet deep. •Medium Dense to Dense Alluvium - Gravelly, silty sand was encountered below the loose alluvium in BH-1 from approximately 30 to 40 feet. In BH-2, medium dense, clean to slightly sand was encountered from approximately 40 to 67 feet, with the upper 5 feet consisting of dense sandy gravel. •Glacial Till - Glacial Till was encountered below the alluvium in both borings, and consisted of unsorted, non-stratified dense to very dense, sandy, gravelly silt to silty, gravelly sand. •Bedrock - The bedrock layer was encountered at a depth of approximately 55 feet in borehole BH-1 at the north bank, but was not encountered within BH-2 at the south bank. This is also a pile foundation bearing strata at the site. The bedrock consisted of fractured basalt, becoming less weathered and stronger with depth. 3.3.2 Ground Water Ground water was observed during drilling in both borings, at depths of approximately 13.5 and 19 feet below the existing ground surface at BH-1 and BH-2, respectively. Because of relatively high permeability of the fill soils and silty sand, it is expected that ground water levels will be reflective of river level. The observed ground water levels during drilling are indicated on the boring logs and on Figure 3. The ground water conditions reported on the exploration logs are for the specific dates and locations indicated and, therefore, may not necessarily be indicative of other times and/or locations. Furthermore, it is anticipated that ground water conditions will vary 2010-100 T200 DR 4 HWA GEOSCIENCES INC. DRAFT February 24, 2015 HWA Project No. 2010-100 T200 in response to other factors such as rainfall, time of year, local subsurface conditions, and other factors. 4.CONCLUSIONS AND RECOMMENDATIONS Geotechnical recommendations are provided below for bridge seismic design criteria, bridge foundations with steel pipe piling, ground improvement to minimize potential liquefaction damages during a design earthquake, slope stability, riprap removal and replacement, and potential construction vibration during ground improvement. 4.1 SEISMIC DESIGN 4.1.1 General Based on the LRFD Bridge Design Specifications (AASHTO, 2013), potential secondary effects of earthquakes on the proposed bridge include soil liquefaction, lateral spreading, seismically- induced settlement, or ground faulting. The following sections provide additional discussions and recommendations pertaining to these seismic issues for use in design of the bridge. 4.1.2 Regional Seismicity The seismicity of northwest Washington is not as well understood as other areas of western North America. Reasons for this include: (1) incomplete old historical earthquake records; (2) deep and relatively young glacial deposits and dense vegetation which obscure surface expression of faults (Hall and Othberg, 1974); and (3) the distribution of recorded seismic epicenters is scattered and does not define mappable fault zones (Gower, et al., 1985). Historical records exist, however, of strong earthquakes with local Modified Mercalli Intensities up to VIII (indicative of structural damage such as cracked walls and fallen chimneys). Since the 1850's, 28 earthquakes of Magnitude 5 (Richter Scale) and greater have reportedly occurred in the eastern Puget Sound and north-central Cascades region. Five events may have exceeded Magnitude 6.0. Researchers consider the North Cascades earthquake of 1872, centered near Lake Chelan, the strongest (Magnitude 7.4) historical earthquake in the region. Earthquakes of Magnitude 7.2 occurred in central Vancouver Island in 1918 and 1946. The most significant recent event, the Nisqually Earthquake, occurred on February 28, 2001, near Olympia and had a magnitude of 6.8. Other significant historical earthquakes in the region include a 1949 event near Olympia (Magnitude 7.2), and a 1965 event centered between Seattle and Tacoma (Magnitude 6.5). These latter three were intraplate Benioff Zone earthquakes, occurring at a depth of about 30 miles within the descending subducted oceanic plate. Potential sources of earthquakes that may be significant to the site include: (1) the Cascadia subduction zone, along which the Juan de Fuca oceanic plate is being thrust under the North 2010-100 T200 DR 5 HWA GEOSCIENCES INC. DRAFT February 24, 2015 HWA Project No. 2010-100 T200 American plate; and (2) shallow crustal faults that may generate earthquakes in the site vicinity (McCrumb, et al., 1989). The latest subduction zone earthquake in the Pacific Northwest had been determined from Japanese tsunami records to have occurred in 1700, and recent offshore sedimentological research has indicated that the entire length of the subduction zone slipped at once, which would result in an earthquake of around Magnitude 9.0. 4.1.3 Seismic Considerations Earthquake loading for the proposed Black River bridge structure was developed in accordance with Section 3.4 of the AASHTO Guide Specifications for LRFD Bridge Design, 2013. For seismic analysis, the Site Class is required to be established and is determined based on the average soil properties in the upper 100 feet below the ground surface. Based on our explorations and understanding of site geology, it is our opinion that the proposed alignment is underlain by soils classifying as Site Class D. Table 1 presents recommended seismic coefficients for use with the general procedure described in the AASHTO, 2013, which is based upon a design event with a 7 percent probability of exceedance in 75 years (equal to a return period of 1,033 years). Ground motions for the site are based on probabilistic earthquake hazard mapping efforts including those conducted by the United States Geological Survey. Accordingly, a Seismic Design Category D, as given by AASHTO, 2013, should be used. Table 1. Seismic Coefficients for Evaluation Using AASHTO Specifications Site Class Peak Ground Acceleration PGA, (g) Spectral Bedrock Acceleration at 0.2 sec Ss, (g) Spectral Bedrock Acceleration at 1.0 sec S1, (g) Site Amplification Coefficients Design Acceleration Coefficient As, (g) Fpga Fa Fv D 0.446 0.993 0.331 1.05 1.1 1.74 0.470 4.1.4 Soil Liquefaction Liquefaction occurs when saturated and relatively cohesionless soil deposits such as silts, sands, and fine gravels temporarily lose strength as a result of earthquake shaking. Primary factors controlling the development of liquefaction include intensity and duration of strong ground motion, characteristics of subsurface soils, in-situ stress conditions and the depth to ground water. Potential effects of soil liquefaction include temporary loss of bearing capacity and lateral soil resistance, and liquefaction-induced settlement and deformations, with concomitant potential impacts on the proposed bridge and embankment fills. Based on the saturated loose nature of the alluvium noted below fill in BH-1 and BH-2, liquefaction shall be a design consideration for this project. 2010-100 T200 DR 6 HWA GEOSCIENCES INC. DRAFT February 24, 2015 HWA Project No. 2010-100 T200 Based on the Seed and Idriss (1971), and Ishihara and Yoshimine (1992), liquefaction for the loose alluvium/fill layer, 20 feet thick, below the road crust fill will liquefy during PGA=0.446g and a Mw=7.5 earthquake. 4.1.5 Ground Fault Hazard The Seattle and Tacoma Faults are probably the most serious earthquake threat to the populous Seattle–Tacoma area. Black River Bridge site is located between these faults. A 2005 study of bridge vulnerability estimated that a magnitude 6.7 earthquake on the Seattle Fault would damage approximately 80 bridges in the Seattle–Tacoma area, whereas a magnitude 9.0 subduction event would damage only around 87 bridges in all of Western Washington. The same study also found that with failure of just six bridges (the minimum damage from a Benioff M 6.5 event) there could be at least $3 billion lost in business revenue alone. Subsequent retrofitting by the Washington Department of Transportation and the City of Seattle would likely reduce damage to key bridges. 4.2 SLOPE STABILITY EVALUATIONS The proposed pedestrian bridge abutments are to be constructed above the top of the river bank slopes. The stability of these slopes was evaluated using limit-equilibrium methods utilizing the computer program SLIDE 5.0 (Rocscience, 2010). Limit equilibrium methods consider force (or moment) equilibrium along potential failure surfaces. Results are provided in terms of a factor of safety, which is computed as the ratio of the summation of the resisting forces to the summation of the driving forces. Where the factor of safety is less than 1.0, instability is predicted. With limit equilibrium, the shear strength available is assumed to mobilize at the same rate at all points along the failure surface. As a result, the factor of safety is constant over the entire failure surface. 4.2.1 Static Slope Stability Analyses The static factors of safety calculated along the Geologic Profile A-A’, Figure 3, was evaluated with Spencer’s method, Janbu’s Simplified method, and Bishop’s Simplified method with the observed condition at the site currently. The factor of safety of the slope at the southern abutment, under static loading, is approximately 1.3 and for the northern abutment is approximately 1.1, as shown on Figures C-1 and C-4 of Appendix C, respectively. This analyses indicates that the factor of safety is slightly greater than unity which means that it is marginally stable under the static condition with the current condition of the slopes. 2010-100 T200 DR 7 HWA GEOSCIENCES INC. DRAFT February 24, 2015 HWA Project No. 2010-100 T200 4.2.2 Pseudo-Static Slope Stability Analyses Geologic Profile A-A’ was evaluated using pseudo-static methods to evaluate the response of the slope under earthquake loading prior to the onset of liquefaction. Spencer’s, Janbu’s Simplified, and Bishop’s Simplified methods were used in this evaluation. Pseudo-static slope stability analysis model the anticipated earthquake loading as a constant horizontal force applied to the soil mass. For our analyses, we used a horizontal seismic coefficient of 0.235g, which is one- half of the peak ground acceleration (PGA or As in Table 1). Pre-liquefaction strengths were used for all materials in this analysis. The results of these analyses indicate a factor of safety of approximately 0.65 and for the northern abutment is approximately 0.62, as shown in Figures C-2 and C-5 of Appendix C, respectively. This analyses indicates that slope instability is likely to occur during the design seismic event, prior to the onset of liquefaction. As a factor of safety less the 1.0 was calculated, we expect the existing slopes to undergo lateral spreading upon the onset of liquefaction. 4.2.3 Post Liquefaction Slope Stability Analyses Additional stability analysis were completed on the slopes depicted in Geologic Profile A-A’ to determine the response of the slopes after the onset of liquefaction. The post liquefaction residual shear strengths for the liquefiable soils were used to model the anticipated loss of shear strength during a seismic event. The results of these analyses indicate a factor of safety of approximately 0.31 and 0.22, as shown in Figures C-3 and C-6 of Appendix C, respectively. As a factor of safety less the 1.0 was calculated, we expect the existing slopes to undergo lateral spreading upon the onset of liquefaction. 4.2.4 Lateral Spreading and Sliding Lateral spreading occurs cyclically when the horizontal ground accelerations combine with gravity to create driving forces which temporarily exceed the available strength of the soil mass. This is a type of failure known as cyclic mobility. The result of a lateral spreading failure is horizontal movement of the partially liquefied soils and any overlying crust of non-liquefied soils. We would expect displacements associated with lateral spreading to be on the order of several feet. Bartlett and Youd (1992) used a large data base of lateral spreading case histories and developed an empirical formula. According to the research, we calculated a yield acceleration (ay=0.2g) by means of a trial and error method for the existing bank slope (2H:1V) and Newmark’s sliding block slope stability analyses. When an earthquake magnitude Mw=7 occurs, the estimated lateral spreading ranges from about 24 to 134 inches depending upon assumed epicenter distances, 60 km (Tacoma Fault) and 6 km (Seattle Fault) away, respectively. Although the results vary widely, the analyses demonstrate that large lateral spreading is a possibility. 2010-100 T200 DR 8 HWA GEOSCIENCES INC. DRAFT February 24, 2015 HWA Project No. 2010-100 T200 A summary of the anticipated factor of safety for global stability at the abutments are provided below in Table 2. Table 2. Global Stability Analyses Results Without GIT Factor of Safety South Side North Side Static 1.3 1.1 Pseudo-Static 0.65 0.62 Post Liquefaction 0.31 0.22 To mitigate these liquefiable soil conditions, we recommend that the proposed bridge be founded on driven piles that extend into the dense glacial till or bedrock at depth, and the strength of the slopes be increased by in-situ ground improvement techniques (GIT); namely vibrocompaction or stone columns. 4.2.5 GLOBAL STABILITY AFTER GROUND IMPROVEMENT Static Slope Stability Analyses The static factors of safety calculated along the Geologic Profile A-A’ were evaluated with Spencer’s method, Janbu’s Simplified method, and Bishop’s Simplified method assuming ground improvement was performed. The factor of safety of the slope at the southern abutment, under static loading assuming GIT, is approximately 1.5 and for the northern abutment is approximately 1.4, as shown on Figures C-7 and C-9 of Appendix C, respectively. This analyses indicates that the factor of safety slightly increased after the application of GIT. Pseudo-Static Slope Stability Analyses Geologic Profile A-A’ was evaluated using pseudo-static methods to evaluate the response of the slope under earthquake loading prior to the onset of liquefaction after the application of GIT. Spencer’s, Janbu’s Simplified, and Bishop’s Simplified methods were used in this evaluation. Pseudo-static slope stability analysis model the anticipated earthquake loading as a constant horizontal force applied to the soil mass. For our analyses, we used a horizontal seismic coefficient of 0.235g, which is one-half of the peak ground acceleration (PGA). Pre-liquefaction strengths were used for all materials in this analysis. 2010-100 T200 DR 9 HWA GEOSCIENCES INC. DRAFT February 24, 2015 HWA Project No. 2010-100 T200 The results of these analyses indicate a factor of safety of approximately 0.84 and for the northern abutment is approximately 0.66, as shown in Figures C-8 and C-10 of Appendix C, respectively. This indicates that slope instability is likely during a seismic event, prior to the onset of liquefaction. As a factor of safety less than 1.0 was calculated, we expect the GIT-treated slopes to undergo minor lateral spreading (non-catastrophic) upon the onset of liquefaction. The summary of the stability analyses is summarized in Table 3, below. Table 3. Global Stability Analyses Results after GIT Factor of Safety South Side North Side Static After GIT 1.5 1.4 Pseudo-Static After GIT 0.84 0.66 4.3 Ground Improvement Techniques (GIT) The bridge foundations should be designed to withstand liquefaction-induced lateral and down- drag loading as well as liquefaction-induced flow slide loading. To mitigate liquefaction conditions and densify the loose sand layer noted below the fill, vibrocompaction (VC) or stone columns (SC), which is sometimes called vibro-replacement, are considered. The process of VC consists of first boring a hole by using air or water jetting with vibrating probe into the granular soils to the required improvement depth. Densification of the loose sand/silt is achieved by excitement applied to the soil at depth. To be effective, the soil should not have more than 15 percent fines. At BH-1, the soil has too much fines for this method to be effective. Also, the water jetting which will be necessary for the operation may not be acceptable due to environmental impacts. We recommend SC treatment for this site due to consideration of silt contents of the formation. The stone column method (SC) is a dry method (without injecting water) by which vertical columns are made of compacted aggregate extending through a deposit of loose soil, and result in increased shear resistance of the slope and relief of pore-water pressure during the design earthquake event. SCs are installed with a deep stone feed tube and with vibratory action of the probe forcing the aggregate radially into the loose soil zones, compacting the stone as well as any granular zones formed in the surrounding soil. Typical diameters of stone columns are 2 to 4 feet. The stone columns will provide dissipation of excess pore pressure during strong shaking and the treated soil layer will not liquefy. 2010-100 T200 DR 10 HWA GEOSCIENCES INC. DRAFT February 24, 2015 HWA Project No. 2010-100 T200 As indicated in the previous sections of slope stability analyses, SC will not completely eliminate the slope instability problem during the design earthquake event, but it will prevent liquefaction of the loose alluvium layer, and thereby reduce lateral spreading (Bohn and Lambert, 2013). Consequently, SC will significantly improve the safety of the pedestrians on the bridge during the design earthquake event. At this site, the Medium Dense to Dense Fill at the surface is thick ranging from 8 to 18 feet, at the north and south bank, respectively. The stone column feed tube may not be able to penetrate the Fill layer to reach the Loose Alluvium layer which needs to be treated. Therefore, we recommend that the surface crust (Fill) be predrilled to insert the stone column tube. The existing slope is armored with riprap stones which shall be removed also prior to inserting the stone column tube. The cost associated with predrilling, removal and restoration of riprap on the slopes shall be included for estimating the cost of the project. The areas of stone column treatment at both river banks shall include a 80-foot wide centered on the bridge centerline. The first row of treatment shall start from 5 feet away from the shoreline, with subsequent rows progressing up slope to 20 feet beyond the pile cap. The stone column spacing shall be 8 feet on center with a triangular pattern. The contractor shall allow the ground a sufficient time for dissipating excess pore pressures due to vibratory SC operations. If sufficient time is not allowed for pore pressure dissipation, slides can occur during SC treatment. It is the contractor’s responsibility to ensure slides do not happen. This treatment shall occur in the dry summer months to take advantage of low water level. The treatment depths shall be down to EL 0 and EL -10 at the north and south banks, respectively. Loose Alluvium thickness to receive SC is about 20 feet, to depths of approximately 30 feet (north side) to 40 feet (south side). 4.4 Ground Improvement Verification Tests After the SC treatment, the soils between stone columns shall be tested with 3 test borings at each side of the river to verify the degrees of densification in terms of relative density, Dr (%). The minimum density shall be greater than 50 percent after stone column treatment. The verification test results shall be documented by means of Standard Penetration Test (SPT) values and the geotechnical engineer of record shall evaluate the soil density improvement. 4.5 BRIDGE FOUNDATIONS Very dense glacial soils or bedrock was encountered below the alluvium in our borings. We recommend driven steel pipe piles (14-inch diameter) with closed ends be used to support the proposed pedestrian bridge. The slope stability analyses as shown on Appendix C indicate that even after the treatment with SC, the slope stability is still not satisfactory under the design level earthquake. However, it is 2010-100 T200 DR 11 HWA GEOSCIENCES INC. DRAFT February 24, 2015 HWA Project No. 2010-100 T200 our opinion that shallow foundations shall not be used and, instead, pile foundations should be used for the bridge. This is to render a minimum support to safeguard pedestrians on the bridge during the design earthquake event. Pile foundations should be driven prior to the SC treatment to avoid interferences with stone column locations. When the pile foundations combined with stone columns are used for this bridge support, the probability of catastrophic bridge failure and human life fatalities will be significantly small. 4.5.1 Axial Loading Foundation design requirements provided by Parametrix are presented in Table 4. Table 4. Foundation Requirements for North and South Abutments Span Type Approximate Ground Elevation (feet) Maximum Scour Elevation (feet) Preliminary Axial Loads (kips) Service I, kips Strength I, kips Extreme I, kips Concrete 30 NA 229 317 209 Steel Truss 30 NA 135 200 91 4.5.2 Down Drag Loading During and after the ground is liquefied, the soils above the liquefied soil layer (loose Sand zone in our case) will tend to drag down the pile by mobilizing skin frictional forces. Such loading is called Down Drag and it can be calculated for during and after the design earthquake event. The fill and the loose sand layer will drag down as the ground settles. The residual angle of 5 degrees was assumed to act on the pile shaft. Assuming 14-inch steel pipe piles, the down drag loading including a 1.25 load factor is calculated to be 52 kips. When the area is treated with ground improvement, the down drag loading shall be ignored. This loading can be ignored, if SC treatment is applied. 4.6 Axial Resistance for Pile Design. The nominal end-bearing and skin frictional shaft resistances are calculated and presented in Table 5. The nominal capacity is sometimes called ‘Ultimate Capacity’. 2010-100 T200 DR 12 HWA GEOSCIENCES INC. DRAFT February 24, 2015 HWA Project No. 2010-100 T200 Table 5. Nominal Axial Resistance for 14 inch Diameter Steel Pipe Piles Nominal Axial resistances, SPT, Meyerhof method North Bank, kips South Bank, kips Resistance factor, ϕstat Rp, end bearing resistance, kips 1099 1282 0.3 Rs, side frictional resistance, kips 57 110 0.3 Nominal Resistance Rn=Rp + Rs 1156 1393 0.3 Resistance factor for the service and extreme cases is unity (1). 4.6.1 Uplift Resistance During and After Earthquake The net uplift resistances are tabulated in Table 6. Frictional resistance from Fill and Loose Alluvium layers during the design earthquake is ignored, because the magnitude of the uplift skin friction is relatively small. Table 6. Net Uplift Capacities, 14 inch Steel Pipe Pile North Bank, kips South Bank, kips Resistance factor, ϕup Net Uplift resistances, Rn 208 330 0.25 4.6.2 Summary of Resistance Factors All capacities tabulated above for single piles are assumed to be positioned with their spacings greater than 3 diameters. If pile spacings are less than 3 diameters, the nominal capacities will need to be reduced to account for group interaction effects. 2010-100 T200 DR 13 HWA GEOSCIENCES INC. DRAFT February 24, 2015 HWA Project No. 2010-100 T200 For the strength limit state design, resistance factors of 0.3 for skin frictional resistance and 0.3 for tip resistance, are recommended. For uplift, the resistance factor of 0.25 should be used. For the extreme and service limit states, we recommend resistance factors of 1.0 and 1.0 for skin resistance and tip resistance, respectively. Our calculations indicate that for the service loads indicated, total pile settlements will be less than one percent (0.5%) of the pile diameter. An acceptable service load settlement (e.g., 1 inch) will be used for the design. Once the piles are driven to driving refusal and the tips have reached the bearing layers (till or bedrock), the settlement of the piles will be very close to the compression of structural element, i.e., the pile shaft elastic compression at this site. 4.7 Laterally Loaded Driven Piles We understand that lateral loads on the shafts will be evaluated using the computer program LPILE (Ensoft, 2002). This program is based on the p-y method (Reese 1984), which was originally developed for slender piles that tend to bend and deflect when subjected to lateral loads and bending moments. We recommend the design parameters presented in Table 7 for lateral analyses with LPILE. Because the loose sandy silts and silty sands are considered susceptible to soil liquefaction, design parameters for static and cyclic loading conditions (during and post-liquefaction) are presented in the table below. Table 7. Recommended Parameters (N & S banks) for Use in LPILE Analyses For Static and Seismic Pre-Liquefaction Soil Type Effective Unit Weight γ’ Friction Angle φ (degrees) Apparent Cohesion c (psf) Modulus of Horiz. Subgrade Reaction, k ε50 (%) pcf degrees psf pci Top Fill 130 36 0 125 -- Loose Sandy Silt 48 28 0 20 MD Sand 60 36 0 60 -- Glacial Till or Bedrock 80 45 0 4000 -- 2010-100 T200 DR 14 HWA GEOSCIENCES INC. DRAFT February 24, 2015 HWA Project No. 2010-100 T200 For During Liquefaction Soil Type Effective Unit Weight γ’ Friction Angle φ (degrees) Apparent Cohesion c (psf) Modulus of Horiz. Subgrade Reaction, k ε50 (%) pcf degrees psf pci Top Fill 130 36 0 125 -- Loose Sandy Silt 48 0 0 0 MD Sand 60 36 0 125 -- Glacial Till or Bedrock 80 45 0 4000 -- For Post-Liquefaction Soil Type Effective Unit Weight γ’ Friction Angle φ Apparent Cohesion c Modulus of Horiz. Subgrade Reaction, k ε50 (%) pcf degrees psf pci Top Fill 130 36 0 125 -- Loose Sandy Silt 48 5** residual 0 2** residual MD to D Sand 60 34 0 125 -- Glacial Till or Bedrock 80 45 0 4000 -- **Brandenberg et al (2007), herein adopted 10% p-multiplier, mp for post liquefaction, (10% of Static k), also per FHWA-NHI-10-016, May 2010, pp 12-61and WSDOT GDM Fig 6-16. For Static and Seismic Conditions after Stone Column Treatment Soil Type Effective Unit Weight γ’ Friction Angle φ (degrees) Apparent Cohesion c (psf) Modulus of Horiz. Subgrade Reaction, k ε50 (%) pcf degrees psf pci Top Fill 130 36 0 125 -- Loose Sandy Silt 58 32 0 55 2010-100 T200 DR 15 HWA GEOSCIENCES INC. DRAFT February 24, 2015 HWA Project No. 2010-100 T200 MD Sand 60 36 0 60 -- Glacial Till or Bedrock 80 45 0 4000 -- 4.8 DRIVEN STEEL PIPE PILE INSTALLATION AND CONSTRUCTION CONSIDERATIONS Piles shall be driven with a minimum hammer energy 35,000 to 45,000 ft-lb rated energy. The tip shall be driven down to glacial till or bedrock; at BH-1 EL -35’ and at BH-2, EL -60’. The pile driving shall be witnessed and inspected by the geotechnical engineer of the record. The contractor shall submit the pile driving hammer Model Numbers with its specifications two week prior to the initiation of driving. Piles should be driven to a required minimum penetration so that they bear in dense to very dense soil, and to the penetration resistance required to achieve a net bearing capacity equal to twice the allowable load, as determined by the Wave Equation analysis of pile driving. All piles shall be driven to meet practical refusal, i.e., typically about 10 blows per inch. This analysis should be performed after the pile lengths, pile-driving hammer, cushion, and pile capblock have been selected by the contractor. 4.9 BRIDGE ABUTMENTS AND WING WALLS 4.9.1 Lateral Earth Pressures - Static Condition Lateral earth pressures used for design of bridge abutments under static loading conditions should be equivalent to that generated by a fluid weighing 55 pcf, assuming tops of the abutments are restrained from lateral movement. An equivalent fluid unit weight of 35 pcf should be utilized if the tops are free to rotate (active case). The above recommendations assume properly compacted, well-drained granular fill adjacent to the abutments. Traffic surcharge loads should also be included in the abutment design. Lateral loads at bridge abutments can be resisted by passive resistance of buried structural elements. However, the passive resistance of soil or structural fill above design scour elevation should not be included in design. In this project, scour will not affect the lateral resistance. If the abutment vertical loads are to be carried by deep foundations, frictional resistance along the base of the abutments should not be included in calculating resistance to lateral loads. Passive resistance may be evaluated using an equivalent fluid density of 300 pcf for structural elements cast neat against undisturbed existing Fill (it is considered to be a structural fill) near the ground surface and the upper two feet shall be ignored for the passive resistances. We recommend a passive pressure resistance factor, φep, of 0.45 be used in design for the strength limit state. For the extreme event limit state, the corresponding factor should be 1.0. The passive 2010-100 T200 DR 16 HWA GEOSCIENCES INC. DRAFT DRAFT February 24, 2015 HWA Project No. 2010-100 T200 we recommend the passive resistance shall be ignored for the design, unless a sufficient inspection is achieved to make sure that all soils are compacted at the toe of walls. 4.9.3 Abutment Wall Backfill Abutment wall design and construction should be in accordance with applicable WSDOT Standards. Wall backfill materials should consist of Gravel Backfill for Walls (WSDOT 9- 03.12(2)), or Gravel Borrow (WSDOT 9-03.14), as described in the WSDOT Standard Specifications (WSDOT, 2014). Placement and compaction of fill behind walls shall be in accordance with WSDOT 2-09.3(1) E, with the exception that the compaction standard referenced in Section 2-03.3(14) D should be Modified Proctor, ASTM D 1557. Wall drainage systems should also be designed and constructed in accordance with the WSDOT Standard Specifications. Provisions for permanent control of subsurface water should at a minimum consist of a perforated drain pipe behind and at the base of the wall, embedded in clean, free-draining sand and gravel. The base of the drain pipe should be a minimum of 12 inches below the base of the adjacent ground surface at the toe of the wall. The drain pipe should be graded to direct water away from backfill and subgrade soils and to a suitable outlet. 4.10 SPREAD FOOTING BEARING CAPACITY ON EXISTING DENSE FILL Shallow strip and square footings supporting bridge approach fills on level ground can be designed with the net bearing capacity (qnet) of 5,000 psf and on the sloped ground (2H:1H) 2000 psf with a 3 feet minimum embedment depth. A resistance factor, ϕstat =0.5, shall be applied for the design. The footing settlement under the load will be less than one inch. The minimum depths of the footings should not be less than 18 inches below ground surface. The footing bottom shall be compacted to the densities as specified in Section 4.12. The resistance factor for the extreme and service cases is one. 4.11 SLIDING RESISTANCE ON EXISTING FILL FOR CAST-IN-PLACE CONCRETE FOOTINGS Friction on compacted fill or the existing Fill at the base of the footing shall be 0.4. Resistance Factor ϕτ =0.8 shall be used. The resistance factor for the extreme and service cases is one. 4.12 STRUCTURAL FILL MATERIALS AND COMPACTION In our opinion, the existing fill on site will be suitable for use as structural fill, providing it is isolated of any fine-grained (silt and clay) or organic rich material. In addition, cobbles and boulders should be screened out of native site soil to be re-used as structural fill. If required, imported structural fill should consist of relatively clean, free draining, sand and gravel conforming to the Gravel Borrow specification, Section 9-03.14 (Gravel Borrow) of the 2014 WSDOT Standard Specifications. If earthwork is performed during extended periods of 2010-100 T200 DR 18 HWA GEOSCIENCES INC. DRAFT February 24, 2015 HWA Project No. 2010-100 T200 wet weather or in wet conditions, the structural fill should conform to the recommendations provided in the Wet Weather Earthwork section following. In general, the backfill should be placed in horizontal lifts and compacted to a dense and unyielding condition, and at least 95 percent of its maximum dry density, as determined by test method described in Section 2-03.3(14)D of the 2002 WSDOT Standard Specifications. The thickness of loose lifts should not exceed 8 inches for heavy equipment compactors and 4 inches for hand operated compactors. The procedure to achieve the specified minimum relative compaction depends on the size and type of compaction equipment, the number of passes, thickness of the layer being compacted, and on soil moisture-density properties. We recommend that the appropriate lift thickness, and the adequacy of the subgrade preparation and materials compaction, be evaluated by a representative of the geotechnical consultant during construction. A sufficient number of in-place density tests should be performed as the fill is being placed to determine if the required compaction is being achieved. 4.13 WET WEATHER EARTHWORK The on-site fill is considered moderately moisture sensitive and may be difficult to traverse with construction equipment during periods of wet weather or wet conditions. Furthermore, the near- surface soils may be difficult to compact if their moisture content significantly exceeds the optimum. General recommendations relative to earthwork performed in wet weather or in wet conditions are presented below. •Earthwork should be performed in small areas to minimize exposure to wet weather. Excavation or the removal of unsuitable soil should be followed promptly by the placement and compaction of clean structural fill. The size and type of construction equipment used may have to be limited to prevent soil disturbance. Under some circumstances, it may be necessary to excavate soils with a backhoe to minimize subgrade disturbance that may be caused by equipment traffic. •Material used as structural fill should consist of clean granular soil with less than 5 percent passing the U.S. Standard No. 200 sieve, based on wet sieving the fraction passing the ¾-inch sieve. The fine-grained portion of the structural fill soils should be non-plastic. •The ground surface within the construction area should be graded to promote run-off of surface water and to prevent the ponding of water. 2010-100 T200 DR 19 HWA GEOSCIENCES INC. DRAFT February 24, 2015 HWA Project No. 2010-100 T200 •The ground surface within the construction area should be sealed by a smooth drum vibratory roller, or equivalent, and under no circumstances should soil be left uncompacted and exposed to moisture. •Excavation and placement of structural fill material should be performed under the full-time observation of a representative of the geotechnical engineer, to determine that the work is being accomplished in accordance with the project specifications and the recommendations contained herein. •Bales of straw and/or geotextile silt fences should be strategically located to control erosion and the movement of soil. 4.14 EMBANKMENT SLOPES We recommend that the planned compacted fill slopes or bank slopes be constructed/restored to no steeper than 2H:1V (horizontal:vertical). For fill slopes constructed at 2H:1V or flatter, and comprised of fill soils placed and compacted as structural fill as described above, we anticipate that adequate factors of safety against global failure will be maintained. Measures should be taken to prevent surficial instability and/or erosion of embankment material. This can be accomplished by conscientious compaction of the embankment fills all the way out to the slope face, by maintaining adequate drainage, and planting the disturbed slope face with vegetation as soon as possible after construction. To achieve the specified relative compaction at the slope face, it may be necessary to overbuild the slopes several feet, and then trim back to design finish grade. In our experience, compaction of slope faces by “track-walking” is generally ineffective and is, therefore, not recommended. Even after the SC treatment on the banks, riprap rocks shall be installed from the toe level of the slopes to the design flood level in the river. The riprap rocks removed from the slopes can be re- used. Riprap rocks (18” minus in diameter) meeting WSDOT 9-13 and 9-13.4(2) shall be underlain by a 12 inch layer of 4 inch minus Quarry Spalls, per WSDOT 9-03.6. If ripraps is not allowed by the agencies, bioengineered erosion protection shall be incorporated into the slope restoration, which is beyond our current scope of work. 4.15 SITE DRAINAGE AND EROSION 4.15.1 Surface Water Control Surface runoff can be controlled during construction by careful grading practices. Typically, these include the construction of shallow, upgrade, perimeter ditches or low earthen berms and the use of temporary sumps to collect runoff and prevent water from damaging exposed subgrades. Also, measures should be taken to avoid ponding of surface water during construction. 2010-100 T200 DR 20 HWA GEOSCIENCES INC. DRAFT February 24, 2015 HWA Project No. 2010-100 T200 Permanent control of surface water should be incorporated in the final grading design. Adequate surface gradients and drainage systems should be incorporated into the design such that surface runoff is directed away from structures and pavements and into swales or other controlled drainage devices. 4.15.2 Erosion Control In our opinion, erosion at the site during construction can be minimized by implementing the recommendations presented in Wet Weather Earthwork, Section 4.13, and by judicious use of straw bales, silt fences and plastic sheets. The erosion control devices should be in place and remain in place throughout site preparation and construction. Potential problems associated with erosion may also be minimized by establishing vegetation within disturbed areas immediately following grading operations. Vegetation with deep penetrating roots is the preferred choice, since the roots tend to maintain the surficial stability of slopes by mechanical effects and contribute to the drying of slopes by evapotranspiration. 5.CONDITIONS AND LIMITATIONS We have prepared this report for use by Parametrix, Inc. and King County in design of a portion of this project. The report and any other applicable geotechnical data should be provided in its entirety to prospective contractors for their bidding or estimating purposes, but our report, conclusions and interpretations should not be construed as a warranty of the subsurface conditions. Experience has shown that subsurface soil and ground water conditions can vary significantly over small distances. Inconsistent conditions can occur between explorations and may not be detected by a geotechnical study. If, during future site operations, subsurface conditions are encountered which vary appreciably from those described herein, HWA should be notified for review of the recommendations of this report, and revision of such if necessary. If there is a substantial lapse of time between the submission of this report and the start of construction, or if conditions have changed due to construction operations at or near the site, it is recommended that this report be reviewed to determine the applicability of the conclusions and recommendations considering the changed conditions and time lapse. This report is issued with the understanding that the information and recommendations contained herein will be brought to the attention of the appropriate design team personnel and incorporated into the project plans and specifications, and the necessary steps will be taken to verify that the contractor and subcontractors carry out such recommendations in the field. Within the limitations of scope, schedule and budget, HWA attempted to execute these services in accordance with generally accepted professional principles and practices in the fields of geotechnical engineering and engineering geology in the area at the time the report was prepared. No warranty, express or implied, is made. The scope of our work did not include environmental 2010-100 T200 DR 21 HWA GEOSCIENCES INC. DRAFT February 24, 2015 HWA Project No. 2010-100 T200 assessments or evaluations regarding the presence or absence of wetlands, hazardous substances in the soil, or surface water at this site. This firm does not practice or consult in the field of safety engineering. We do not direct the contractor’s operations, and cannot be responsible for the safety of personnel other than our own on the site. As such, the safety of others is the responsibility of the contractor. The contractor should notify the owner if he considers any of the recommended actions presented herein unsafe.  We appreciate the opportunity to provide geotechnical services on this project. Should you have any questions or comments, or if we may be of further service, please do not hesitate to call. Sincerely, HWA GEOSCIENCES INC. Brad W. Thurber, L.G, L.E.G. Sa H. Hong, P.E. Senior Engineering Geologist Principal Geotechnical Engineer 2010-100 T200 DR 22 HWA GEOSCIENCES INC. DRAFT 6. REFERENCES AASHTO, 2013, LRFD Bridge Design Specifications, Third Edition. Bartlett, S.F., Youd, T.L., 1992, Empirical analysis of horizontal ground displacement generated by liquefaction-induced lateral spread, Tech report NCEER-92-0021. Bohn Cecilia and Lambert Serge, 2013, Case Studies of Stone Columns Improvement in Seismic Areas, 3rd Conference, Maghrebine en Engenierie Geotechnique. Brandenberg et al, 2011, Recommended Design Practice for Pile Foundations in Laterally Spreading Ground, Pacific Earthquake Engineering Research Center. EERI and Washington Military Dept. – Emergency Management Division, 2005, Scenario for a Magnitude 6.7 Earthquake on the Seattle Fault. Ensoft, Inc (2002), Documentation of Computer Program LPILE. Gower, H. D., J.C. Yount and R.S. Crosson, 1985, Seismotectonic Map of the Puget Sound Region, Washington, U. S. Geological Survey, Miscellaneous Investigations Series Map I- 1613. Hall, J.B. and K.L. Othberg, 1974, Thickness of Unconsolidated Sediments, Puget Lowland, Washington, State of Washington, Department of Natural Resources, Division of Geology and Earth Resources. Idriss, I. M., and Boulanger, R. W., 2007, SPT and CPT based relationships for the residual shear strength of liquefied soils, Earthquake Geotechnical Engineering, Proc., 4th International Conf. on Earthq. Geotech. Engineering. Ishihara, K. and Yoshimine, M., 1992, Evaluation of Settlements in Sand Deposits following Liquefaction during Earthquakes, Soils and Foundations, Vol 15, No. 1, pp 29-44. Kramer, S.L., 1996, Geotechnical Earthquake Engineering, Prentice Hall. McCrumb, D., et al., 1989, Tectonics, Seismicity, and Engineering Seismology in Washington, Engineering Geology in Washington, Vol. 1, Washington Division of Geology and Earth Resources Bulletin 78. Seed, H.B., Idriss, I. M. 1971, Simplified procedure for evaluating soil liquefaction potential, J. Soil Mech. Found. Div. 23 DRAFT U.S. Geological Survey, 1996, National Seismic Hazard Mapping Project. Washington State Department of Transportation (WSDOT), 2011, Geotechnical Design Manual (GDM), M 46-03.06. Washington State Department of Transportation (WSDOT), 2014, Standard Specifications for Road, Bridge and Municipal Construction. DRAFT FIGURE NO. PROJECT NO. Approximate Project Site Location NORTH NOT TO SCALE BASE MAP FROM GOOGLE MAPS- DATA MAP © 2015 VICINITY MAP BLACK RIVER BRIDGE LAKE TO SOUND TRAIL RENTON, WASHINGTON 1 2010-100 T200 DRAFT DRAFT DRAFT APPENDIX A FIELD INVESTIGATION DRAFT APPENDIX A FIELD INVESTIGATION Two geotechnical borings were drilled for the proposed Black River Bridge, on November 10, 2014 and January 6, 2015. These borings were designated BH-1 and BH-2, and were drilled at the top of the river banks in the general centerline of the proposed bridge alignment. The borings were drilled to maximum depths ranging from 61 to 86.5 feet below the existing ground surface. The exploration locations were located in the field by taping distances from known site features and plotted. The locations of the borings are indicated on Figures 2 and 3. The borings were drilled by Holocene Drilling, Inc. of Puyallup, Washington, under subcontract to HWA Geosciences Inc. The borings were advanced using a track-mounted, Dietrich D50 drill rig equipped with hollow stem augers. Each of the explorations was completed under the full- time supervision and observation of an HWA geologist. Soil samples were collected at 2.5- to 5-foot intervals using Standard Penetration Test (SPT) methods in general accordance with ASTM D-1586. SPT sampling consisted of using a 2-inch outside diameter, split-spoon sampler driven with a 140-pound drop hammer using a rope and cathead. During the test, a sample is obtained by driving the sampler 18 inches into the soil with the hammer free-falling 30 inches per blow. The number of blows required for each 6 inches of penetration is recorded. The Standard Penetration Resistance ("N-value") of the soil is calculated as the number of blows required for the final 12 inches of penetration. This resistance, or N- value, provides an indication of the relative density of granular soils and the relative consistency of cohesive soils. HWA personnel recorded pertinent information including soil sample depths, stratigraphy, soil engineering characteristics, and ground water occurrence. Soils were classified in general accordance with the classification system described in Figure A-1, which also provides a key to the exploration log symbols. Representative soil samples were taken to our laboratory for further examination. The summary logs of boreholes are presented on Figures A-2 and A-3. The stratigraphic contacts shown on the individual logs represent the approximate boundaries between soil types; actual transitions may be more gradual. Moreover, the soil and ground water conditions depicted are only for the specific locations and dates reported and, therefore, are not necessarily representative of other locations and times. 2010-100 T200 DR A-1 HWA GEOSCIENCES INC. DRAFT A-12010-100-200 Renton, Washington Black River Bridge Lake to Sound Trail SYMBOLS USED ON EXPLORATION LOGS LEGEND OF TERMS AND to 30 over 30 Approximate Undrained Shear Strength (psf) <250 250 - No. 4 Sieve Sand with Fines (appreciable amount of fines) amount of fines) More than 50% Retained on No. 200 Sieve Size Sand and Sandy Soils Clean Gravel (little or no fines) More than 50% of Coarse Fraction Retained on No. 4 Sieve Gravel with SM SC ML MH CH OH RELATIVE DENSITY OR CONSISTENCY VERSUS SPT N-VALUE Very Loose Loose Medium Dense Very Dense Dense N (blows/ft) 0 to 4 4 to 10 10 to 30 30 to 50 over 50 Approximate Relative Density(%) 0 - 15 15 - 35 35 - 65 65 - 85 85 - 100 COHESIVE SOILS Consistency Very Soft Soft Medium Stiff Stiff Very Stiff Hard N (blows/ft) 0 to 2 2 to 4 4 to 8 8 to 15 15 Clean Sand (little or no fines) 50% or More of Coarse Fraction Passing Fine Grained Soils Silt and Clay Liquid Limit Less than 50% 50% or More Passing No. 200 Sieve Size Silt and Clay Liquid Limit 50% or More 500 500 - 1000 1000 - 2000 2000 - 4000 >4000 DensityDensity USCS SOIL CLASSIFICATION SYSTEM Coarse Grained Soils Gravel and Gravelly Soils Highly Organic Soils GROUP DESCRIPTIONS Well-graded GRAVEL Poorly-graded GRAVEL Silty GRAVEL Clayey GRAVEL Well-graded SAND Poorly-graded SAND Silty SAND Clayey SAND SILT Lean CLAY Organic SILT/Organic CLAY Elastic SILT Fat CLAY Organic SILT/Organic CLAY PEAT MAJOR DIVISIONS GW SP CL OL PT GP GM GC SW COHESIONLESS SOILS Fines (appreciable LEGEND 2010-100-200.GPJ 2/20/15 FIGURE:PROJECT NO.: Coarse sand Medium sand SIZE RANGE Larger than 12 in Smaller than No. 200 (0.074mm) Gravel time of drilling) Groundwater Level (measured in well or AL CBR CN Atterberg Limits:LL = Liquid Limit California Bearing Ratio Consolidation Resilient Modulus Photoionization Device Reading Pocket Penetrometer Specific Gravity Triaxial Compression Torvane 3 in to 12 in 3 in to No 4 (4.5mm) No. 4 (4.5 mm) to No. 200 (0.074 mm) COMPONENT DRY Absence of moisture, dusty, dry to the touch. MOIST Damp but no visible water. WET Visible free water, usually soil is below water table. Boulders Cobbles Coarse gravel Fine gravel Sand MOISTURE CONTENT COMPONENT PROPORTIONS Fine sand Silt and Clay 5 - 12% PROPORTION RANGE DESCRIPTIVE TERMS Clean Slightly (Clayey, Silty, Sandy) 30 - 50% Components are arranged in order of increasing quantities. Very (Clayey, Silty, Sandy, Gravelly) 12 - 30%Clayey, Silty, Sandy, Gravelly open hole after water level stabilized) Groundwater Level (measured at 3 in to 3/4 in 3/4 in to No 4 (4.5mm) No. 4 (4.5 mm) to No. 10 (2.0 mm) No. 10 (2.0 mm) to No. 40 (0.42 mm) No. 40 (0.42 mm) to No. 200 (0.074 mm) PL = Plastic Limit DD DS GS K MD MR PID PP SG TC TV Dry Density (pcf) Direct Shear Grain Size Distribution Permeability Approx. Shear Strength (tsf) Percent Fines%F Moisture/Density Relationship (Proctor) Approx. Compressive Strength (tsf) Unconfined CompressionUC (140 lb. hammer with 30 in. drop) Shelby Tube Small Bag Sample Large Bag (Bulk) Sample Core Run Non-standard Penetration Test 2.0" OD Split Spoon (SPT) NOTES: Soil classifications presented on exploration logs are based on visual and laboratory observation. Density/consistency, color, modifier (if any) GROUP NAME, additions to group name (if any), moisture content. Proportion, gradation, and angularity of constituents, additional comments. (GEOLOGIC INTERPRETATION) Please refer to the discussion in the report text as well as the exploration logs for a more complete description of subsurface conditions. Soil descriptions are presented in the following general order: < 5% 3-1/4" OD Split Spoon with Brass Rings (3.0" OD split spoon) TEST SYMBOLS SAMPLE TYPE SYMBOLS GROUNDWATER SYMBOLS COMPONENT DEFINITIONS DRAFT SM ML ML SM SM 6-9-10 9-9-8 4-5-5 2-2-3 0-0-0 1-0-1 0-0-0 1-1-1 1-2-2 3-6-10 6-10-10 GS GS GS GS GS Medium dense, gray, silty to slightly silty, fine SAND, moist. BLocky texture, light brown at surface, trace organics and burnt wood bits. (FILL) Medium dense, dark grayish brown, sandy SILT, moist. Very loose to medium dense, gray, fine sandy SILT, moist to wet. Trace organic bits and layers, some laminar bedding. (ALLUVIUM) Blow counts are weight of hammer only. Sample is wet at tip of sample. Ground water seepage was observed at 13.5 feet below ground surface. Abundant organics in sample. Laminar layers or organics. Lots of heave encountered, 4-5 feet cleaned out of auger. Loose, grayish brown, silty, fine SAND, wet. No recovery of sample. Medium dense, gray, silty, gravelly, fine to coarse SAND, wet. Wood bits and organics observed. (ALLUVIUM) S-1 S-2 S-3 S-4 S-5 S-6 S-7 S-8 S-9 S-10 S-11 BORING 2010-100-200.GPJ 2/20/15 FIGURE:PROJECT NO.:2010-100-200 Renton, Washington Black River Bridge Lake to Sound Trail DRILLING COMPANY: Holocene Drilling DRILLING METHOD: Diedrich D-50 track rig with HSA SAMPLING METHOD: SPT Autohammer LOCATION: See Figure 2 DATE STARTED: 11/10/2014 DATE COMPLETED: 11/10/2014 LOGGED BY: D. ColtranefeetSURFACE ELEVATION: For a proper understanding of the nature of subsurface conditions, this exploration log should be read in conjunction with the text of the geotechnical report. 26.50 0 5 10 15 20 25 30 35 40DEPTH (feet)DEPTH(feet)0 5 10 15 20 25 30 35 40 BH-1 PAGE: 1 of 2(blows/6 inches)A-2GROUNDWATEROTHER TESTSPlastic Limit BORING: and therefore may not necessarily be indicative of other times and/or locations.PEN. RESISTANCENOTE: This log of subsurface conditions applies only at the specified location and on the date indicated DESCRIPTION (140 lb. weight, 30" drop) Blows per foot Liquid LimitSYMBOL0 10 20 30 40 50 0 20 40 60 80 100SAMPLE TYPESAMPLE NUMBERNatural Water ContentUSCS SOIL CLASSWater Content (%) Standard Penetration Test DRAFT SM 9-7-16 12-27-44 26-13-15 12-10-25 50-50/2" GSMedium dense to very dense, light brown, gravelly, silty, fine to coarse SAND, wet. Angular gravel and sands, blocky texture. (GLACIAL TILL) Bluish gray, moderately weak, highly weathered, fractured, basalt. Speckled coloring. (WEATHERED BEDROCK) Bluish gray, moderately strong, moderately weathered, fractured, BASALT. Speckled coloring. (TUKWILA FORMATION) Boring was terminated at 61 feet below surface in bedrock. Ground water seepage was observed at 13.5 feet below ground surface. S-12 S-13 S-14 S-15 S-16 BORING 2010-100-200.GPJ 2/20/15 FIGURE:PROJECT NO.:2010-100-200 Renton, Washington Black River Bridge Lake to Sound Trail DRILLING COMPANY: Holocene Drilling DRILLING METHOD: Diedrich D-50 track rig with HSA SAMPLING METHOD: SPT Autohammer LOCATION: See Figure 2 DATE STARTED: 11/10/2014 DATE COMPLETED: 11/10/2014 LOGGED BY: D. ColtranefeetSURFACE ELEVATION: For a proper understanding of the nature of subsurface conditions, this exploration log should be read in conjunction with the text of the geotechnical report. 26.50 40 45 50 55 60 65 70 75 80DEPTH (feet)DEPTH(feet)40 45 50 55 60 65 70 75 80 BH-1 PAGE: 2 of 2(blows/6 inches)A-2GROUNDWATEROTHER TESTSPlastic Limit BORING: and therefore may not necessarily be indicative of other times and/or locations.PEN. RESISTANCENOTE: This log of subsurface conditions applies only at the specified location and on the date indicated DESCRIPTION (140 lb. weight, 30" drop) Blows per foot Liquid LimitSYMBOL0 10 20 30 40 50 0 20 40 60 80 100SAMPLE TYPESAMPLE NUMBERNatural Water ContentUSCS SOIL CLASSWater Content (%) Standard Penetration Test DRAFT SP SM ML SM SM SP GP SP 20-18-19 9-12-15 4-5-7 5-10-10 3-6-6 5-7-8 2-2-2 30-1-2 1-2-4 3-3-3 4-3-3 Grass at surface. Dense, light brown, slightly silty, slightly gravelly, SAND, moist. Broken gravels and concrete. (FILL) Medium dense, gray, slightly gravelly, very sandy SILT, moist. Wood bits observed. Loose, gray and brown, interbedded fine to medium SAND with SILT layers, moist to wet. Ground water seepage observed at 19.0 feet during drilling. Loose, gray, slightly silty SAND, wet. Initial 6-inch blow count is from chunk of rubber in sampler. Loose, gray, sandy, fine GRAVEL, wet. (ALLUVIUM) Loose, gray, slightly silty, fine to medium SAND, wet. S-1 S-2 S-3 S-4 S-5 S-6 S-7 S-8 S-9 S-10 S-11 BORING 2010-100-200.GPJ 2/20/15 FIGURE:PROJECT NO.:2010-100-200 Renton, Washington Black River Bridge Lake to Sound Trail DRILLING COMPANY: Holocene Drilling DRILLING METHOD: Diedrich D-50 track rig with HSA SAMPLING METHOD: SPT Autohammer LOCATION: See Figure 2 DATE STARTED: 1/6/2015 DATE COMPLETED: 1/6/2015 LOGGED BY: D. ColtranefeetSURFACE ELEVATION: For a proper understanding of the nature of subsurface conditions, this exploration log should be read in conjunction with the text of the geotechnical report. 29.00 0 5 10 15 20 25 30 35 40DEPTH (feet)DEPTH(feet)0 5 10 15 20 25 30 35 40 BH-2 PAGE: 1 of 3(blows/6 inches)A-3GROUNDWATEROTHER TESTSPlastic Limit BORING: and therefore may not necessarily be indicative of other times and/or locations.PEN. RESISTANCENOTE: This log of subsurface conditions applies only at the specified location and on the date indicated DESCRIPTION (140 lb. weight, 30" drop) Blows per foot Liquid LimitSYMBOL0 10 20 30 40 50 0 20 40 60 80 100SAMPLE TYPESAMPLE NUMBERNatural Water ContentUSCS SOIL CLASSWater Content (%) Standard Penetration Test DRAFT GP SP SP SM GM 6-16-19 4-8-11 2-11-10 7-10-11 6-10-13 6-20-19 9-11-20 Dense, gray, slightly silty, sandy GRAVEL, wet. Broken gravels in sampler. Medium dense, black, fine to medium SAND, wet. Bits of wood noted in samples. No sample recovery, shells noted in cuttings. Medium dense, dark gray, slightly silty, fine to medium SAND, wet. Shells observed. Drilling becomes gravelly. Poor recovery; broken gravel. Dense, gray, slightly sandy, gravelly, SILT, wet. Broken gravels in sampler. (GLACIAL TILL) S-12 S-13 S-14 S-15 S-16 S-17 S-18 BORING 2010-100-200.GPJ 2/20/15 FIGURE:PROJECT NO.:2010-100-200 Renton, Washington Black River Bridge Lake to Sound Trail DRILLING COMPANY: Holocene Drilling DRILLING METHOD: Diedrich D-50 track rig with HSA SAMPLING METHOD: SPT Autohammer LOCATION: See Figure 2 DATE STARTED: 1/6/2015 DATE COMPLETED: 1/6/2015 LOGGED BY: D. ColtranefeetSURFACE ELEVATION: For a proper understanding of the nature of subsurface conditions, this exploration log should be read in conjunction with the text of the geotechnical report. 29.00 40 45 50 55 60 65 70 75 80DEPTH (feet)DEPTH(feet)40 45 50 55 60 65 70 75 80 BH-2 PAGE: 2 of 3(blows/6 inches)A-3GROUNDWATEROTHER TESTSPlastic Limit BORING: and therefore may not necessarily be indicative of other times and/or locations.PEN. RESISTANCENOTE: This log of subsurface conditions applies only at the specified location and on the date indicated DESCRIPTION (140 lb. weight, 30" drop) Blows per foot Liquid LimitSYMBOL0 10 20 30 40 50 0 20 40 60 80 100SAMPLE TYPESAMPLE NUMBERNatural Water ContentUSCS SOIL CLASSWater Content (%) Standard Penetration Test DRAFT 12-13-17 50/4" Becomes medium dense, broken gravel in sampler. Very dense, gray, sandy, gravelly SILT, wet. Most likely driven on boulder. Boring terminated at 86.5 feet below ground surface due to refusal. Ground water seepage was observed at 19 feet below ground surface during the exploration. S-19 S-20 BORING 2010-100-200.GPJ 2/20/15 FIGURE:PROJECT NO.:2010-100-200 Renton, Washington Black River Bridge Lake to Sound Trail DRILLING COMPANY: Holocene Drilling DRILLING METHOD: Diedrich D-50 track rig with HSA SAMPLING METHOD: SPT Autohammer LOCATION: See Figure 2 DATE STARTED: 1/6/2015 DATE COMPLETED: 1/6/2015 LOGGED BY: D. ColtranefeetSURFACE ELEVATION: For a proper understanding of the nature of subsurface conditions, this exploration log should be read in conjunction with the text of the geotechnical report. 29.00 80 85 90 95 100 105 110 115 120DEPTH (feet)DEPTH(feet)80 85 90 95 100 105 110 115 120 BH-2 PAGE: 3 of 3(blows/6 inches)A-3GROUNDWATEROTHER TESTSPlastic Limit BORING: and therefore may not necessarily be indicative of other times and/or locations.PEN. RESISTANCENOTE: This log of subsurface conditions applies only at the specified location and on the date indicated DESCRIPTION (140 lb. weight, 30" drop) Blows per foot Liquid LimitSYMBOL0 10 20 30 40 50 0 20 40 60 80 100SAMPLE TYPESAMPLE NUMBERNatural Water ContentUSCS SOIL CLASSWater Content (%) Standard Penetration Test DRAFT APPENDIX B LABORATORY INVESTIGATION DRAFT APPENDIX B LABORATORY TESTING Laboratory tests were performed on selected samples obtained from the borings to characterize relevant engineering and index properties of the site soils. Because of the predominantly coarse- grained nature of the encountered soils, the collected and tested samples should not be considered representative of the existing soils. For the same reason, only a limited number of laboratory tests could be performed on the obtained soil samples. HWA personnel performed laboratory tests in general accordance with appropriate ASTM test methods. We tested selected soil samples to determine moisture content and grain-size distribution. The test procedures and results are briefly discussed below. Moisture Content Laboratory tests were conducted to determine the moisture content of selected soil samples, in general accordance with ASTM D-2216. Test results are indicated at the sampled intervals on the appropriate boring logs in Appendix A. Grain Size Analysis The grain size distributions of selected soil samples were determined in general accordance with ASTM D 422. Grain size distribution curves for the tested samples are presented on Figures B-1 through B-4. 2010-100 T200 DR B-1 HWA GEOSCIENCES INC. DRAFT 0 10 20 30 40 50 60 70 80 90 100 0.0010.010.1110 BH-1 BH-1 BH-1 5.0 - 6.5 7.5 - 9.0 12.5 - 14.0 SILT 3/4" GRAVEL % MC LL PL PI 90 GRAIN SIZE IN MILLIMETERS 0.05 5/8" 70 1.7 0.7 #10 40.4 21.7 15.0 (ML) Dark grayish brown, Sandy SILT (ML) Dark grayish brown, SILT with sand (ML) Gray, SILT with sand #20 Fine Coarse DEPTH (ft)SYMBOL Gravel%Sand%Fines% 30 CLASSIFICATION OF SOIL- ASTM D2487 Group Symbol and Name U.S. STANDARD SIEVE SIZES SAND 10 3"1-1/2"PERCENT FINER BY WEIGHT#4 #200 S-2 S-3 S-5 16 23 36 50 SAMPLE B-1 0.00050.005 CLAY #100 0.5 50 Medium Fine 57.9 77.6 85.0 3/8" 5 Coarse #60#40 PARTICLE-SIZE ANALYSIS OF SOILS METHOD ASTM D422 2010-100-200PROJECT NO.: HWAGRSZ 2010-100-200.GPJ 2/20/15 FIGURE: Lake to Sound Trail Black River Bridge Renton, Washington DRAFT 0 10 20 30 40 50 60 70 80 90 100 0.0010.010.1110 BH-1 BH-1 BH-1 17.5 - 19.0 25.0 - 26.5 40.0 - 41.5 SILT 3/4" GRAVEL % MC LL PL PI 90 GRAIN SIZE IN MILLIMETERS 0.05 5/8" 70 15.3 #10 20.4 71.8 39.4 (ML) Gray, SILT with sand and organics (SM) Grayish brown, Silty SAND (SM) Yellowish brown, Silty SAND with gravel #20 Fine Coarse DEPTH (ft)SYMBOL Gravel%Sand%Fines% 30 CLASSIFICATION OF SOIL- ASTM D2487 Group Symbol and Name U.S. STANDARD SIEVE SIZES SAND 10 3"1-1/2"PERCENT FINER BY WEIGHT#4 #200 S-7 S-9 S-12 56 38 25 50 SAMPLE B-2 0.00050.005 CLAY #100 0.5 50 Medium Fine 79.6 28.2 45.3 3/8" 5 Coarse #60#40 PARTICLE-SIZE ANALYSIS OF SOILS METHOD ASTM D422 2010-100-200PROJECT NO.: HWAGRSZ 2010-100-200.GPJ 2/20/15 FIGURE: Lake to Sound Trail Black River Bridge Renton, Washington DRAFT 0 10 20 30 40 50 60 70 80 90 100 0.0010.010.1110 BH-2 BH-2 BH-2 7.5 - 9.0 15.0 - 16.5 45.0 - 46.5 SILT 3/4" GRAVEL % MC LL PL PI 90 GRAIN SIZE IN MILLIMETERS 0.05 5/8" 70 1.2 3.2 5.7 #10 46.8 44.2 89.9 (ML) Dark grayish brown, sandy SILT (ML) Dark grayish brown, sandy SILT (SP) Black, Poorly graded SAND #20 Fine Coarse DEPTH (ft)SYMBOL Gravel%Sand%Fines% 30 CLASSIFICATION OF SOIL- ASTM D2487 Group Symbol and Name U.S. STANDARD SIEVE SIZES SAND 10 3"1-1/2"PERCENT FINER BY WEIGHT#4 #200 S-3 S-6 S-13 19 18 29 50 SAMPLE B-3 0.00050.005 CLAY #100 0.5 50 Medium Fine 52.0 52.7 4.3 3/8" 5 Coarse #60#40 PARTICLE-SIZE ANALYSIS OF SOILS METHOD ASTM D422 2010-100-200PROJECT NO.: HWAGRSZ 2010-100-200.GPJ 2/20/15 FIGURE: Lake to Sound Trail Black River Bridge Renton, Washington DRAFT 0 10 20 30 40 50 60 70 80 90 100 0.0010.010.1110 BH-2 BH-2 BH-2 60.0 - 61.5 75.0 - 76.5 86.0 - 86.5 SILT 3/4" GRAVEL % MC LL PL PI 90 GRAIN SIZE IN MILLIMETERS 0.05 5/8" 70 0.9 27.7 12.3 #10 92.0 22.7 23.5 (SP-SM) Dark gray, Poorly graded SAND with silt (GM) Gray, Silty GRAVEL with sand (ML) Gray, SILT with sand #20 Fine Coarse DEPTH (ft)SYMBOL Gravel%Sand%Fines% 30 CLASSIFICATION OF SOIL- ASTM D2487 Group Symbol and Name U.S. STANDARD SIEVE SIZES SAND 10 3"1-1/2"PERCENT FINER BY WEIGHT#4 #200 S-16 S-18 S-20 23 21 21 50 SAMPLE B-4 0.00050.005 CLAY #100 0.5 50 Medium Fine 7.1 49.6 64.2 3/8" 5 Coarse #60#40 PARTICLE-SIZE ANALYSIS OF SOILS METHOD ASTM D422 2010-100-200PROJECT NO.: HWAGRSZ 2010-100-200.GPJ 2/20/15 FIGURE: Lake to Sound Trail Black River Bridge Renton, Washington DRAFT APPENDIX C SLOPE STABILITY ANALYSES, COMPUTER CALCULATION RESULTS DRAFT STATIC STABILITY: SOUTH ABUTMENT BLACK RIVER BRIDGE LAKE TO SOUND TRAIL RENTON, WA C-1 2010-100-21 FIGURE NO. PROJECT NO. DRAFT SEISMIC STABILITY: SOUTH ABUTMENT (DESIGN EVENT) BLACK RIVER BRIDGE LAKE TO SOUND TRAIL RENTON, WA C-2 2010-100-21 FIGURE NO. PROJECT NO. DRAFT POST LIQUEFACTION STABILITY: SOUTH ABUTMENT (DESIGN EVENT) BLACK RIVER BRIDGE LAKE TO SOUND TRAIL RENTON, WA C-3 2010-100-21 FIGURE NO. PROJECT NO. DRAFT STATIC STABILITY: NORTH ABUTMENT BLACK RIVER BRIDGE LAKE TO SOUND TRAIL RENTON, WA C-4 2010-100-21 FIGURE NO. PROJECT NO. DRAFT SEISMIC STABILITY: NORTH ABUTMENT (DESIGN EVENT) BLACK RIVER BRIDGE LAKE TO SOUND TRAIL RENTON, WA C-5 2010-100-21 FIGURE NO. PROJECT NO. DRAFT POST LIQUEFACTION STABILITY: NORTH ABUTMENT (DESIGN EVENT) BLACK RIVER BRIDGE LAKE TO SOUND TRAIL RENTON, WA C-6 2010-100-21 FIGURE NO. PROJECT NO. DRAFT STATIC STABILITY AFTER GROUND IMPROVEMENTS: SOUTH ABUTMENT BLACK RIVER BRIDGE LAKE TO SOUND TRAIL RENTON, WA C-7 2010-100-21 FIGURE NO. PROJECT NO. DRAFT PSEUDO-STATIC STABILITY AFTER GROUND IMPROVEMENTS: SOUTH ABUTMENT BLACK RIVER BRIDGE LAKE TO SOUND TRAIL RENTON, WA C-8 2010-100-21 FIGURE NO. PROJECT NO. DRAFT STATIC STABILITY AFTER GROUND IMPROVEMENTS: NORTH ABUTMENT BLACK RIVER BRIDGE LAKE TO SOUND TRAIL RENTON, WA C-9 2010-100-21 FIGURE NO. PROJECT NO. DRAFT PSEUDO-STATIC STABILITY AFTER GROUND IMPROVEMENTS: NORTH ABUTMENT BLACK RIVER BRIDGE LAKE TO SOUND TRAIL RENTON, WA C-10 2010-100-21 FIGURE NO. PROJECT NO. DRAFT